The Restoration of St. Helena s Church in .NET

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The Restoration of St. Helena s Church
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FIGURE 7-5 The sidewalls appeared to have settled, although diagonal settlement cracks were not visible.
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there would be a horizontal component of the beam reaction normal to the bearing surface. The masonry wall is 27 inches thick below the balcony sill beam. If the beam was installed green, shrinkage would have occurred across the grain as much as a quarter of an inch. A small amount of shrinkage at the sill could act as a notch, causing a hinged effect in the masonry wall. The eccentricity of the notch, combined with rotation at the top of the wall due to truss defection, would be suf cient to cause the amount of rotation that was observed in the wall. This, I determined, was the principal mechanism causing the balcony wings to rotate away from the building. It was signi cant that the horizontal crack is located at the thinnest portion of the masonry wall at the top of the sill beam.
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FIGURE 7-6 The 10 10 sills embedded in the masonry sidewalls were somewhat decayed.
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One of the two balcony beams observed had a sloped bearing surface at upper end rather than a horizontally cut seat. The stucco at the hinge in the exterior wall had been patched several times. This indicated that the movement of the wall had been progressive over time. In the computer, the north and south sidewalls were modeled as an integral part of the building cross-section. We utilized a low modulus of elasticity of 350,000 psi for the brick masonry. Assumed allowable design values for the masonry are as follows: Compressive strength, 56 psi Bearing, 100 psi Shear, 9 psi The 12-foot spacing of the building cross-section modeled in the computer included one roof truss, a bay of balcony framing, and the masonry sidewalls with window openings. Overall building stability in the cross-section appeared tenuous, except for the balcony acting as a deep horizontal beam fastened to the endwalls. In the analysis we included the horizontal contribution of the balcony as a spring support based on the stiffness of the plywood sheathing on the balcony steps. There was evidence that the balcony, with its considerable plywood sheathing,
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The Restoration of St. Helena s Church
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was performing well in that there was no cracking of the plaster ceiling under the balcony. Of course, all of the lateral loads applied to the sidewalls needed to be resolved through the balcony sheathing, roof sheathing, and ceiling framing into the endwalls of the nave. I argued that through jacking, the exterior wall could be forced back into alignment. In order to do this, both the balcony and roof loads would have to be temporarily relieved and the sidewall temporarily braced. All mortar joints that had opened over time and have been lled, in the vicinity of the fold, would have to be raked out in order to reverse the alignment. The gap above the balcony sill would have to be lled and the end of the balcony beam at cut, blocked, or tied in where the sloped bearing conditions exist. The balcony sill beam would have to be replaced where deteriorated, with a dry timber that would not shrink over time.
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The tower was rebuilt in 1940 in accordance with architectural drawings by architect Albert Simmons of Charleston. The 1940 architectural tracings (drawings) for the tower were stored in the Fireproof Building in Charleston in at les. The tower as built included wood siding on diagonal sheathing on wood studs attached to a steel framing system.
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FIGURE 7-7 The bell tower was reframed in 1940 with wood on a steel frame.
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The steel frame provided great rigidity to the steeple up to the transition level where the framing changes to timber. The bell support system appeared to be much older than the 1940s work. The new tower structure was apparently designed to accommodate the geometry of the bell and its framing system. The lantern and spire sections were framed in timber in a con guration similar to St. Michael s in Charleston. The west side of the tower rested on two cast-in-place concrete blocks supported by the original west wall. The east side of the tower rested on a 30-inch-deep steel beam that spanned the masonry opening between the narthex and the nave. Critical to the bell tower were loading conditions that included northsouth wind, which would tend to place unbalanced loads onto the supporting wall or the W30 beam. Wind in the east-west direction would impose the greatest total load on the 30-inch-deep steel beam. In the computer analysis of the tower, we applied a 120 mph wind in accordance with ASCE-7, except that a shape factor of 0.8 was used for the spire and lantern levels. The bell and bell frame were included as a 1,500-pound load. The analysis was performed in the north-south and east-west directions. The calculations indicated that the steel structure was rigid and that most of the movement in the spire is a result of the open lantern level. Diagonal sheathing was simulated in the timber-framed portion of the steeple to provide a reasonable amount of rigidity. A theoretical de ection of 4 inches at the top of the spire was calculated with these assumptions. The 30-inch-deep steel beam, steel tower legs, and the steel cross-bracing appeared to be adequate in size. Critical to the performance of the steeple in a 120 mph wind load was its anchorage to the top of the outside wall and the anchorage of the 30-inch-deep steel beam to the inside wall on either side of the choir loft. The 22,000 pounds of resistance to hold the steeple legs down could be achieved by assuring that the tie-down points contained a minimum of 150 cubic feet of concrete, or 225 cubic feet of brick masonry. The bearing pressures at all support points were adequate as long as each location has 240 square inches of contact against the brick. The uplift capacity of the steeple could be enhanced by rebuilding the beam and tower bearing points with a compatible but harder brick rather than concrete and facilitating a tie that engages at least 225 cubic feet of masonry.
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